設計中學綜合

2.7m4.5m,3.3m，如下圖所示：ABCDDCDCBA截面寬度取600×1/2=250mm，可得梁的截面初步定為b×h=250*600。樓板取120mm，100mm,250×400。ib）l）ic1 0.02×20＝0.4 0.12×25＝3 0.015×16＝0.24 3.98 0.25×0.6×25＝3.75KN/m 2×（0.6－0.1）×0.02×17＝0.34KN/m 0.25×0.4×25＝2.5KN/m 2×（0.4－0.1）×0.02×17＝0.204KN/mg6AB!=g6cd1=4.09KN/m 0.025×20＝0.5 0.12×25＝3 0.015×16＝0.24 3.24 4.09 2.704gAB!=1=18.75KN/mgBC1=2.704KN/mgAB2=2=19.44KN/mgBC2=8.748KN/m 0.25×0.6×6×25=22.5 1.2×6×0.25×19=34.2 99.46 0.25×0.6×6×25=22.5KN 0.02×（0.6-0.1）×2×6×17=2.04KN連系梁傳來屋面自重0.5×（6+6-2.7）×1.35×3.98=24.980.5×6×3.98×3=35.82頂層點集中荷 85.34 22.5 2.04 2×1.2×1.8×0.45=1.944 0.25×(6-2×1.2-0.5)×1.8×19=26.51 137.98 22.5 2.04 -2.1×0.8×(5.24-0.2)×=-16.93 24.75 1.907

0.5×6×1.5×3.24=14.58p6AB=p6CD=1.5×6=9KN/mp6BC=1.5×2.7=4.05KN/mp6A=p6D=3×3×1.5=13.5pAB=pCD=1.5×6=9KN/mpBC=2×2.7=5.4KN/mpA=pD=3×3×1.5=13.5風壓標準值計算 W=βz. H＜30m，可取 65432111平對橫向柱網(wǎng)影響即可采用剪力法計算水平作用力為求基底剪力，頂層總重力荷載1203 =750.12 六層總重力荷載代表值18089 3.98×81.1×14.7=4744 4＋11.4×14]+0.3×0.25×（2.7-0.6）×14×25＝1467KN ＝6106 0.8×2.1×49×5.24＝431 五層總重力荷載代表值17207 3.24×81.1×14.7＝3862 11.4×14]+0.3×0.25×（2.7-0.6）×14×25＝1467KN ＝6106 0.8×2.1×49×5.24＝431 17207 3.24×81.1×14.7＝3862 11.4×14]+0.3×0.25×（2.7-0.6）×14×25＝1467KN ＝6106 0.8×2.1×49×5.24＝431 三層總重力荷載代表值17207 3.24×81.1×14.7＝3362 11.4×14]+0.3×0.25×（2.7-0.6）×14×25＝1467KN ＝6106 0.8×2.1×49×5.24＝431 二層總重力荷載代表值17207 3.24×81.1×14.7＝3362 11.4×14]+0.3×0.25×（2.7-0.6）×14×25＝1467KN ＝6106 0.8×2.1×49×5.24＝431 一層總重力荷載代表值19812 3.24×81.1×14.7＝3362 4＋11.4×14]+0.3×0.25×（2.7-0.6）×14×25＝1467KN ＝8510 0.8×2.1×49×5.24＝431 11 即=0.52設防烈度8度，由《抗震規(guī)范》查得水平 αmax=0.16按近震Ⅰ類場地取Tg＝0.2，則 FEK＝α1Geq=α10.85GE=0.067力為FEK=FEK/14=6.1×103/14=439KN因T1<1.4Tg,故可不考慮頂部附加作用系數(shù)每榀框架所承受的作用力Gi3Fi65Ge=Gn1

T=1.7ψ 注：uTGi作為水平荷載而算得的結uT按以下計算：（△u）i=VGi/∑D注：∑Dij為第i（△u）i為第i（△u）k為第kGi為G6和Ge之和。654321T1=1.7ψT主，故可用底部剪力法計算水平作用，即： 8аmax=0.16 3、結構總的水平作用標準值FEk因1.4Tg=1.4×0.3=0.42s<T1=0.526s，所以應考慮頂部附加水平作用。頂部附加作用系數(shù)作用下各樓層水平層間剪力Vi為654321∑水平作用下框架結構的層間位移（△u）i和頂點位移ui分別按下列公（△u）i=Vi/∑Du）i654321四、水平作用下框架內(nèi)力計算：Vij=DijViMbij=Vij*yh ）邊柱kyMb）Mu）654321）中kYMbMu6543212、梁端彎矩、剪力及柱軸力分別按以下計算：Mlb=ilb（Mbi+1，jMui，j）/（ilbirb）Mrb=irb（Mbi+1，jMui，j）/（ilbirb）Vb=（Mlb+Mrb）/lNi=∑（Vlb-Vrbblbbl邊柱中柱654321--- ----------=62.42KN=8.78AN頂=132.95+62.42=195.37柱重：0.65×0.65×3.6×25=38.02KNN底=N頂+38.02=233.39KNBN頂=173.39+64.42+8.78=246.59654321BC：VB=VC=3.3×1.2/2=2.16KNA：N頂N底=34.56+14.4=51.84B：N頂N底=49.68+14.4+2.16=69.126543211.2SGk+1.4SQk，1.35SGk+1.00.8 br +V126AM6VB左M3V0B右MV5AM2V9B左M2V9BM右V4AM4V9B左M7V9B右MV3AM4V9B左M7V9B右MV2AM7V9B左M4V9B右MV1AM0V9B左M5V9B右MV若VA-（2q1+q2）al/2>0x>al，則若VA≤0，則2）VA=-x1AB左震：MA=270.47/0.75=360.63KN·mMB=-367.14/0.75=-489.52VA=-（MA-MB）/l+q1l/2+（1-=-30.22則MmaxMmax=1.3MEk-=375.72 MA=-406.29/0.75=-541.72KN·mMB=226.48/0.75=301.97KN·mVA=-（MA-MB）/l+q1l/2+（1-=205.03x= =241.98γREMmax=0.75×241.98=181.48層次123跨層次456跨 V=γ[η（M q1=1.2×4.5=5.4KN/mq2=1.2×（19.26+0.5×6）=27.43KN/mVGb=5.4×6/2+27.43×4.8/2=85.27KNln=6-0.65=5.35mb Ml=18.54/0.75=24.72bbMr=-132.97/0.75=-177.29b V=γ[η（M =91.71b Ml=147.21/0.75=196.28bbMr=6.4/0.75=8.53b V=γ[η（M =92.09q1=1.2×2.5=3.0KN/mq2=1.2×（9.63+0.5*3.6）=13.72KN/mVGb=3.0×2.4/2+13.72×1.2/2=11.83KNln=2.4-0.65=1.75m Mlb=Mrb=36.72/0.75=48.96V=γRe[ηvb（Mlb+Mrb）/ln+V=59.23 Mlb=Mrb=57.21/0.75=76.28V=γRe[ηvb（Mlb+Mrb）/ln+V=87.33 q1=1.2×8.46=10.15q2=1.2×（14.22+0.5×6）=21.38KN/mVGb=10.15×6/2+21.38×4.8/2=87.85KN q1=1.2×2.5=3.0q2=1.2×（7.11+0.5×3.6）=10.69KN/m橫向框架AMM12N6M32N746692M4N9488245M002N53425MN745484M889N3335515M220N53377773M445N1111137M8842N31137502M888N9002260M82253N90048831M447N88812023M9910N88838246kMM12N6M78867N92072M8N122945M88201N9539M9901N17514M11731N922366M00511N1222083M33841N69943M44026N699552M55328N677770M44283N6779921M33673N688117M99956N688339[uN]=N/Ac[uN]=N/Ac43[uNN/Acηc1.2。654321 N=fA=0.15×14.3×6502/103=902.26 6543216SEk=（M上+M下（Mb+Ml/hn12654321（Mb+Ml/hn12654321跨間：Mmax=281.79KN·m支座A：Mmax=406.29KN·m支座Bl：M=367.14KN·m調(diào)整后剪力：V=182.70KN1/2 II（f=f，=310N/mm2 fb，h，（ =1948.90KN·m>994.06 cm α=M/（fb，h cm cm A=ξfb，h cm s2Ф25、2Ф22，A=1742mm2。sy cm y cm 受壓鋼筋，As，=1742mm2As，即αs=[M-fy,As，(h0- s2Ф25、2Ф22，A=2724mm2。Bl上部：s s 2Ф25、2Ф22，A=2724mm2。ρ=2724/300/565=1.6%>ρ=0.3%，A，s ρsv=Mξ計算As實配As實配1A跨間2A跨間3A4A跨間5A跨間6Ab0A柱654321B柱654321柱截面寬度：b=600混凝土抗壓強度設計值：fcm=14.3N/mm2柱端剪力計算值：Vc=192.89/0.75=257.19柱軸力NN=1746.62/0.75=2328.83（1）最不利組合一（調(diào)整后：Mmax=328.12KN·m，N=1746.62=290.92附加偏心矩ea取20mm和偏心方向截面尺寸的1/30兩者中的較大值，即又l0/h<15，取得η=1+=645.62As,=As=[Ne-ξ(1-0.5ξ)fcmbh02]/fy,/(h0-=[1746.62×103×645.62-0.264×(1-0.5×0.264)×14.3×600(2)：Nmax=2198.06KNM=-此組內(nèi)力是非組合情況，且無水平荷載效應，故不必進行調(diào)整。e0=M/N=17.93×106/(2198.06×103)長細比l0/h=4700/600=6.71>5，故應考慮偏心矩增大系數(shù)η。又l0/h<15，取得η=1+As合力點之間的距離ξ=（N-ξbfcmbh0）/[(Ne-0.45fcmbh02)/(0.8-ξb)/(h0-as,)+fcmbh0N>ξbfcmbh0及Ne>0.43fcmbh02.故1An=0.353考慮作用組合的柱軸向壓力設計值故取N=1746.62KN注：Asvi、li為第is<10d=200mm,4Ф10@150，柱的配筋圖如下圖柱123—二—二—二Mηξ大小大小大小4Ф10@1504Ф10@1504Ф10@150柱456—二—二—二Mηξ大小大小大大44444柱123—二—二—二Mηξ大小大小大小44Ф10@15044Ф10@15044Ф10@150柱456—二—二—二Mηξ大小大小大小4Ф8@1504444以第1層點為例，由節(jié)點兩側梁的受彎承載力計算節(jié)點核芯區(qū)的剪力×3.3+0.6×4.5=4.48ΣMb為節(jié)點左右梁端逆時針或順時針方向組合彎矩設計值之和，即=1998.67節(jié)點核芯區(qū)截面的抗震驗算是按箍筋和混凝同抗剪考慮的，設計時，1.5。bj=bc=600mm, 4Ф10@100，則 123點點點58772363521111@@@@+fA(h-a,)/s]/γyvsvj 18941456點點點71+fA(h-a,)/s]/γyvsvj 9對于樓板，根據(jù)塑性理論，l02/l01<3性鉸設計。（1恒載：g=1.2×3.24=3.89KN/m2q+g=3.89+2.6=5.84（2 活載：q=1.3×（2.0+0.2）=2.86恒載：g=1.2×3.98=4.78KN/m2q+g=2.86+4.78=7.64KN/m2、內(nèi)跨：l0=lc-b（lc為軸線長、bh/l01=120/6000=1/60≥1/50，符合構造要求。l1/4取m2=аm1,а=1/n2=1/4=0.25 （其中n為長短跨比值）2M1u+2M2u+M1uI+M1uII+M2uI+M2uII=Pul012（3l02-l01）/12 =5650mm =5650=m1（5.65-== 2 MI=M 2 2M+2M+MI+MII+MI+MII=Pl2（3 m1=4.77KN·mm2=0.25*4.78=1.19KN·m mI=0，mII=（-2）*1.40=-2.80KN·m 按塑性鉸計算彎矩表（KN·m（1-5層樓面）1626345I67I89I0I01626345I67I89I0I0配筋φ8@200，實有As=251配筋φ8@130，實有As=387配筋φ6@180，實有As=157配筋φ6@200，實有As=141一、二、樓梯板計算： 1m p=6.0×1.2+2.0×1.4=10KN/mln=3.3n n=10h0=100-20=80 syA=M/（rfh）=10×10 sys選Φ10@110,A=714s三、平臺板計算： p=3.49×1.2+1.4×2.0=6.99KN/m0 0=1.57h0=100-20=80As=M/（rsfyh0）=1.57×106/0.985/210/80=94.88選Φ6@140,As=202四、平臺梁計算： 荷載分項系數(shù)rG=1.2 p=14.89×1.2+5.4×1.4=25.43KN/m =28.61Lbf，=b+5hf，=250+5×70=600mmh0=400-35=365mmTAs=M/（rsfyh0）=28.61×106/0.985/210/365=378.92Φ18,實有As=509mm2則 贅述?，F(xiàn)考慮罕遇水平作用下框架層間彈塑性位移計算。一、 s s 實配fs1A2A3A4A5A6A二、注：fcmk為混凝土彎曲抗壓強度標準值sA柱23456B柱23456三、A6：因Mbu<Mcu，223.00KN·m<719.57KN·m，所以，Mc6u=Mbu=223.00KN·mA5：因Mbu<ΣMcu，304.68KN·m<719.57+719.74KN·m，所以，Mc6l=Mbu×k6/（k5+k6）=304.68/2=152.34KN·mMcu6l=719.57KN·m152.34KN·mMc5u=Mbu×k5/（k5+k6）=304.68/2=152.34KN·mMcu5u=719.74KN·m152.34A4：Mc5l=203.38Mc4u=203.38A3：Mc4l=254.34Mc3u=254.34A2：Mc3l=273.34Mc2u=273.34A1：Mc2l=627.46×1.0/（1.0+1.03）=309.09KN·mMc1u=627.46×1.03/（1.0+1.03）=318.37KN·mA0：Mc1l=ΣMcu1l=783.38B5：Mc6l=224.22Mc5u=224.22B4：Mc5l=308.41Mc4u=308.41B3：Mc4l=401.10Mc3u=401.10B2：Mc3l=429.60Mc2u=429.60B1：Mc2l=（627.46+355.64）×1.0/（1.0+1.03）=484.29KN·mMc1u=（627.46+355.64）×1.03/（1.0+1.03）=498.82KN·mB0：Mc1l=ΣMcu1l=783.55四、Vyij理論依據(jù)：第i層第j根柱的受剪承載力計算為Vyij=（Mciju+Vy6A=（223.00+152.34）/（3.3-0.6）=125.11KNVy5A=（152.34+203.38）/（3.3-0.6）=151.57KNVy4A=（203.38+254.34）/（3.3-0.6）=152.57KNVy3A=（254.34+273.34）/（3.3-0.6）=175.89KNVy2A=（273.34+309.09）/（3.3-0.6）=194.14KNVy1A=（318.37+783.38）/（4.5-0.6/2）=250.40KNVy6B=（307.02+224.22）/（3.3-0.6）=177.08KNVy5B=（224.22+308.41）/（3.3-0.6）=177.54KNVy4B=（308.41+401.10）/（3.3-0.6）=236.50KNVy3B=（401.10+429.60）/（3.30.6）=276.90KNVy2B=（429.60+484.29）/（3.3-0.6）=304.63KNVy1B=（498.82+783.55）/（4.5-0.6/2）=291.45KN五、Vyi則Vy6=（Vy6A+Vy6B）×2=（125.11+177.08）×2=604.38KNVy5=（151.57+177.54）×2=658.22KNVy4=（152.57+236.50）×2=778.14KNVy3=（175.89+276.90）×2=905.58KNVy2=（194.14+304.63）×2=997.54KNVy1=（250.40+291.45）×2=1083.70KN六、罕遇下彈性樓層剪力Ve的計算8度水平影響系數(shù)最大值αmax=0.9，此時可用0.9/0.16的比值剩以多遇作用下層間彈性剪力Vi求出VeVi的計算結果在第四部分已經(jīng)算出，Ve6=1241.908×0.9/0.16=6985.73KNVe5=2240.645×0.9/0.16=12603.63KNVe4=3052.662×0.9/0.16=17171.22KNVe3=3673.615×0.9/0.16=